HOV means Hoogwaardig Openbaar Vervoer in Dutch which roughly translates to High Quality Public Transport.
Why a new bridge?
The town of Spijkenisse and the island of Voorne-Putten can be reached by only a few connections to the mainland. Apart from the Spijkenisserbrug (Spijkenisse Bridge), the Hartel bridge is an important link in the roads network. The Hartel bridge is also an important connection for bicycles and mopeds to the harbour of Rotterdam north of the Hartel Canal.
Currently the Hartel Bridge consists of 3 lanes of which 1 is used for both directions depending on the traffic. The rest of the bridge is used as bicycle lane. The bridge is due for major maintenance and at the same time the lay-out will be changed to 2 times 2 lanes. This means that there will be no space left for bicycles. Solutions for this were found in a study by Royal Haskoning and the province of Zuid Holland has decided to go for a separate bicycle bridge in between the existing bridge and the Hartel Storm Surge Barrier.
Figure 1 Muiderbrug before renovation
The Muider Bridge is a 300 metre-long steel bridge that carries the A1 motorway to Amsterdam over the Amsterdam Rhine Canal. This bridge was constructed between 1969 and 1970. The orthotropic bridge deck consists of longitudinal steel ribs and transverse cross beams running perpendicular to the length, which are supported by three main beams. The main beams were constructed as steel box girders (figure 2). The bridge crosses the Amsterdam Rhine Canal at an angle of 60 degrees and has a ‘sloping’ end and sloping supporting lines (which lie parallel to the Amsterdam Rhine Canal). In order to allow for the construction of rush-hour lanes, the bridge needed to be widened. Furthermore, damage especially to the orthotropic road surface was observed at various locations as a result of fatigue. Because of the future increase in not only traffic volume but also the axle loads, it became necessary to increase the fatigue resistance and therefore the lifespan of the orthotropic road surface. The selected solution, which was developed by RWS (Dutch Ministry of Transport and Public Works) in conjunction with market parties, is the replacement of the bitumen deck surface with a layer of steel fibre-reinforced, high strength concrete (HSC). The layer of HSC significantly reduces the local stress variations in the orthotropic bridge deck. However, in order to be able to absorb the increase in permanent loads (high strength concrete) and mobile loads (widening and increased axle loads), the static load capacity of the bridge had to be increased first. These works were performed by the joint venture CFE NL / Victor Buyck Steel Construction. The HSC is being installed in a follow-up project.
Figure 2 Original section of the bridge
In the pre-project stage, before the project was brought onto the market, the Building Services department of the Dutch Ministry of Transport and Public Works (now the ‘Dienst Infrastructuur’) carried out a study into possible alternatives, resulting in the following options:
– Increasing the stiffness of the main girders or adding additional main girders
– Reducing the sagging moment through the use of external pre-stressing
– Modifying the bending moments through jacking the (mid) pier supports
– Creating an additional support point in the main span by means of a cable-stayed structure
– Converting the bridge into a suspension bridge.
After the criteria of effectiveness, cost price and available clearance under the bridge deck had been analysed, it was concluded that a combination of jacking the pier supports, together with an additional support point by means of a cable-stayed structure, would be the best solution.
2. strengthening with the cable-stayed structures
For the selected cable-stayed solution, an additional analysis was carried out between: four pylons (both sides of the canal and both sides of the bridge), two pylons (both sides of the canal but on the same side of the bridge) and two pylons placed asymmetrically against the bridge (opposite sides of the canal and opposite sides of the bridge). Ultimately the latter option was selected, as the parallelogram-shaped bridge could then be supported by two cable-stayed systems with identical (short) spans to the middle of the bridge. The skew angle is 60o.
An additional benefit of this cable-stayed solution was the fact that everything could be installed while traffic was still using the bridge. This was a major advantage for a bridge located in one of the most important arteries of the Dutch motorway network.
Through the cable-stayed structures, an additional spring support was created in the middle of the main span. This central support consists of a cross beam which lies perpendicular to the longitudinal length and connects the three main girders. The cross beam extends out past the bridge deck and forms the lifting points for the two cable-stayed structures. Each cable-stayed structure consists of a concrete pylon, a steel compression beam, three front stay cables, three back stay cables and an anchorage. The tilted pylon (in both directions) crosses with the compression beam, which is installed in the pylon using a dowel and pressed against it using pre-stressing bars. The anchorage consists of a massive concrete foundation slab, which is anchored underground using tension piles. Underneath this slab is a cellar, where the anchors of the pre-stress cables haven been installed.
Figure 3 lay-out of cable-stayed structure
3. Jacking the existing bridge and replacing the bearings
The cross beam consists of a steel box girder of the same height as the existing main girders. Installing the two parts of the cross beam between the main girders increased the dead weight of the bridge deck by such an extent that the bottom of the bridge deck just touched the bridge clearance for ships passing under the bridge on the Amsterdam Rhine Canal. It was therefore decided that both overhangs of the cross beam could only be installed once the bridge had been jacked.
At the end of the deck, the bridge rests on one bearing per main girder. On either side of each abutment’s bearing, an anchorage cable stresses the bridge girder against the bearing.
At the intermediate supporting points, the main girders rest on the pier support walls with two bearings per girder. The bearings showed a lot of wear and did not provide sufficient load capacity for the new bridge. Therefore they had to be replaced, which was done during jacking of the bridge.
The jacking process was split into three phases: one per abutment and, lastly, a phase in which the jacking took place on the two banks. On the site of the pier supports, the bridge had to be jacked to a minimum height of 150 mm and a maximum height of 182 mm. Likewise, the central main girder had to be jacked to 165 mm. The jacking process had to be carried out simultaneously for both pier supports. The difference in jacked height between two adjacent main girders on the same bank also was not allowed to exceed 5 mm during the entire jacking process. For this reason, it was decided that this jacking process would be completely carried out using a computer-aided lifting system. In this system, all jacks are controlled by a computer system, whereby the oil feed into the different jacks is wholly determined on the basis of continuous measurement of the jack strokes that are being realised at that moment.
Figure 4 Jacks at pier supports and strengthening plates
The cantilever beams next to the main girders, under which the bearings are located, had strengthening plates against which the jacks could be installed. Verifying the existing strength of these beams revealed that, these jack plates were only just able to support the dead weight of the bridge during its assembly and therefore had to be strengthened as the jacking process would now be carried out while traffic was using the bridge.
Based on the calculations strengthening plates had to be fixed to the lower and upper flanges. In addition, stiffeners had to be installed inside these beams to correctly transfer the local forces from the jacks to the beams.
As the jacks were installed at the ends of the concrete pier supports, the pier supports also had to be verified. Here, an analysis was carried out to determine whether the existing reinforcement could absorb the possible primary, secondary and angle splitting forces. On the basis of the research carried out into the existing concrete strength, in which the ageing effect of the concrete strength was reconfirmed and in which Fritz Leonhardt’s theories in “Vorlesungen über Massivbau”  were applied, it could be proven that the existing concrete pier supports could absorb the forces from the jacking process without additional strengthening.
4.1 Final design phase
To keep the disruption to the traffic on the A1 motorway to a minimum, the pylons were designed as pre-fabricated structures. Only the lower 14 metres of the 70 metre-tall pylons were constructed on the foundation block with in situ concrete. The remaining pylon height was constructed in 11 prefabricated segments, each approximately 5.10 m in height. The stay cable anchorage can be found in the seventh segment (with its upper side at a height of 50 metres), to which the front and rear stay cables are attached. The top four segments were only included for architectural reasons.
The pylon was designed with a vertical tilt in two directions. In the longitudinal direction of the bridge, the angle between the horizontal plane and the axis of the pylon is approximately 75°. In the direction crossing the bridge deck, this angle is 84.5°. In its final state, the pylon can be considered as a free cantilevered column structure that is affected by stay cable forces at a fixed distance from the column base.
The pylon has been modelled with a rigid support at the base of the structural model. Because the dimensions of the pylon were determined on the basis of applied (stay cable) forces, the interaction between the pylon, stay cables and the existing bridge were not introduced in these calculations. This reduced the pylon to a statically determined, limited cantilevered structure. For the calculation of the occurring action-effects of the pylon as a whole, the stiffness of the rigid support is irrelevant, meaning it is assumed to be indefinitely stiff. If, on the other hand, the calculation of the deformations is considered, the stiffness of the foundation as well as the interaction with the cable-stayed structure and the existing bridge did have to be taken into account thoroughly.
An important part of the design is the horizontal joint between the segments. In connection with the large shear stresses occurring in the horizontal joint, which are the result of the decomposition of the stay cable forces to a horizontal plane, at an earlier stage, a tooth-shaped connection in the joints between the segments was considered. As this was very difficult to realise, a design with flat horizontal joints was chosen eventually.
In each joint and for every load combination, the action-effects Nrep, Mx;rep and My;rep were determined. In this way, the stresses in each corner of the cross section could be determined. As a result, a geometry of pre-stressed cables was deducted using an iterative process. After a suitable course of the pre-tensioning cables over the height of the pylon was found, for each joint, the joint filling and/or the adjacent concrete surface still had to be verified to ensure the maximum occurring concrete compressive stress (ULS) would not exceed its maximum allowed value. Furthermore, verification was also required to ascertain whether the joint filling and/or the adjacent concrete surface were able to resist the occurring shear stresses. On the one hand, these shear stresses were caused by shear forces from the stay cable forces and, on the other hand, by the horizontal components of the pre-tensioning forces in the tilted pre-tensioning cables. Finally, these shear stresses were also increased by a torsion moment, because of the non-convergence of the shear centre and centre of gravity from the sections considered and the torsion moments from the stay cable forces.
The calculated shear stresses that occurred were checked for:
failure of the compression diagonals
|τ2 = 0.2knkθf’b > τd||(1)|
permissible shear stresses
|τu = kbfb + ks.(Fpw – Nd) / Ab > τd||(2)|
Resulting from the above mentioned design calculations, the course of the pre-tensioning cables was determined as detailed in the 3D-picture above.
4.2 Design of pylon assembly phase
At the start of the design of the assembly phase, final agreements were made regarding the placement of the prefabricated pylon sections. If the wind speed exceeded force 6 on the Beaufort scale, the works had to be stopped. In addition, the delivery speed of the segments in the crane was limited to 0.033 m/s vertically and 0 m/s horizontally (i.e. only delivery from above). As a result, the impact load on the free cantilevered pylon was reduced as much as possible.
In the joint, three flat jacks were used. They were pumped full of water up to a height of approx. 25 mm. Then, the new segment was placed on these three jacks. Two previously embedded pilot pins were used for this – these are steel bolts with a point in the top side of segment ‘n’, which fit into sockets embedded in the underside of segment ‘n+1’. Next, two pre-tensioning cables were installed and partially stressed. The horizontal components of the pre-tensioning force were absorbed by the pilot pins. Following on from this, the joint filling was applied between both segments. After segment ‘n+2’ was placed on (again) three jacks, the strength of the joint (hardening) between segments ‘n’ and ‘n+1’ was measured. Once the joint was strong enough, the flat jacks could be removed from this joint to enable the joint to take over the weight of the segments above. Only then could two additional pre-tensioning cables, which had been installed in the tensioning ridges of segment ‘n+2’, be (partially) tensioned.
This methodology was maintained up to and including the installation of segment 6. No pre-tensioning was applied between segments 6 and 7. Although this seems very strange to many people, the pre-tensioning of this joint is provided by the stay cables themselves because the stay cable anchorage is located just in segment 7. Owing to a well-balanced tensioning protocol, the normal stresses in this joint remained sufficiently high so as to allow the shear stresses that occurred.
For the architectural top segments (segments 8 to 11 inclusive), a different methodology was used. These segments do not have any pre-tensioning, but are all connected by GEWI rods. The segments are placed on three rubber bearings. To avoid instability, they were connected to the segment below with two temporary external anchoring connections. These are M30 threaded rods that are tightened in sockets, which are welded to embedded, anchored plates on both sides of the joint. Two temporary connections were installed for each joint, as the specifications stipulated that a double safety measure was required. Following this, GEWI rods were inserted into ducts and connected to the GEWI rods of the segment below using screw couplings. Next, the joint and the ducts around the GEWIs were grouted. Following sufficient hardening, the next segment could be installed.
From the perspective of the double safety measure, the installation of segments 1 to 6 inclusive was constantly based on the possibility that one of the two pre-stressed strands that had been stressed last could fail.
The entire execution process of the pylon constantly brought with it changing loading cases and load combinations for each joint. Accordingly, calculations were carried out either for the shear and or normal stresses in the joint, or the compressive forces in the jacks/rubber bearings and the tensile forces in the temporary external anchoring connections. Because of the speed of the realisation process, during the assembly phases the joints were monitored on the basis of their minimum present strength, for which the criteria were formed based on, on one hand, no tensile stress and, on the other hand, the monitoring of the permissible compressive stresses and shear stresses. From these calculations it was concluded that the filling material should, at least, possess the concrete quality B15 (f’ck = 15 N/mm2) as per the Dutch classifications, before the flat jacks could be lowered.
5. Pre-tensioning procedure
Royal Haskoning carried out the detailed design of the pylon and the anchorage block. The detailed design of the compression beam and stay cables was carried out by Greisch. Greisch elaborated the pre-tensioning procedure, for which Royal Haskoning checked the forces resulting from the proposed procedure in relation to the pylon resistance and, if desired, adjusted or confirmed it. This iterative procedure eventually led to a detailed sequence of steps, where, on the one hand, certain strands were pre-tensioned and, on the other, strands were added.
At the request of Freyssinet, the number of relocations of the jacks during the set process was reduced to a minimum. However, the pylon structure proved to be rather sensitive at a relatively advanced stage of installation and a large number of small sub-steps were necessary. The pre-tensioning in the dowel pre-tensioning bars and in the anchorage cables had to be increased at certain times.
In view of the seemingly constant presence of traffic, wind and temperature influences, the pre-tensioning procedure could not be determined on the basis of the strand forces to be realised, but rather elongations that were to be put into effect had to be assumed in each set step. This was based on the initial strand lengths between the ‘working points’ of each stay cable, these being the straight lengths between the fronts of the anchorage plates on both stay cable ends. As the pylon (with cable anchorage) and the compression beam were put in place in the same weekend as the first procedure steps, the pre-tensioning protocol was determined entirely on the basis of theoretical strand lengths. A check and/or adaptation took place in the first six hours after the placement and measurement of the anchorage positions in the pylons.
6. Monitoring and execution
The isotension method of Freyssinet was used during the pre-tensioning of the strands. This meant that the master strand was given the required extension, which resulted in a definite stress in this strand. Then another strand was stressed until the moment that the stress in that strand was the same as the stress in the master strand at that moment (and thus lower than the initial stress in the master strand). This methodology was maintained until all the strands in the step had been stressed. After finishing a discrete protocol step, which can consist of several sub-steps that can be carried out simultaneously, the force in each of the 16 stay cables was measured. The measured forces were then tested with a monitoring tool. This is a spreadsheet, where theoretically expected forces are reproduced in relation to zones of alertness.
As well as monitoring the stay cable forces, prisms were also secured to the ends of the compression beams, the cross beam, the top segment of the pylon (segment 7, the upper segments were put in place after the pre-tensioning process), the four corners of the pylon foundation and the anchorage foundation block. These prisms were measured every 15 minutes during the phase before contact between the compression beams and cross beam. At the beginning of the 12-hour traffic diversion, this was increased to every 2 minutes. This meant that all measured displacements could be compared with the projected displacements in the calculations.
After set step 31 was completed, a short night-time traffic ban was introduced, in which all stay cable forces and all measurements of the prisms were combined for a final calibration of the end positions. After determining and executing this final position, all strands could be cut off and the cover sockets placed and injected on the stay anchorages. After removing the temporary scaffolding on both pylons, the top segments were put in place, which meant that the transformation from girder bridge to cable-stayed bridge had been successfully completed.
My special thanks goes to Hans Mortier from CFE for his input concerning the execution of the works, Wouter Smit and Ming Yan Liu from Royal Haskoning, Marijn Laethem from Greisch and the people from CFE, RWS and VBSC for our good cooperation during this project. Special thoughts also go to Felix Scholten, our lead engineer during this project, who passed away unexpectedly before completion of the project.
 Fritz Leonhardt and Eduard Mönning, “Vorlesungen Über Massivbau”, 3rd edition, Springer-Verlag, 1977.